J911 Studies Peer Reviewers

[Newtons Bit]

I've never really been generally impressed by the folks at the "Journal" of 911 Studies, but my latest encounter with one of their peer reviewers, Tony Szamboti, on this forum leads me to believe that they're not only suffering from group think, but also egregiously incompetent on issues they proclaim to be experts on. Mr. Szamboti posited this question to a group of non-engineers:

What does the slenderness ratio of a structural steel column need to be to be in the inelastic buckling range?

What were the slenderness ratios of the central core columns at the collapse initiation sites of the 98th floor in the North Tower and 82nd floor in the South Tower?

I saw this question and jumped into the discussion, posting:

Inelastic buckling occurs for all slenderness ratios under the Euler limit. That's 4.71 * SQRT(E/Fy). Of course extremely stout members won't buckle inelastically, however none of the columns in the upper floors of the WTC were that stout.

Do you have any clue as to what you're talking about? I recommend picking up an AISC Manual of Steel Construction and see exactly how steel is designed these days. We're not in the 1940's, we know how steel fails now. Maybe you should update you knowledge to modern information.

Mr. Szamboti's replied:

How did I know you would come on.

You used an effective length factor of 1.0 in your letter to Gordon Ross, which is for a pinned connection, when you should have used .5 to .65 for fixed both ends connections for the tower columns. The 1.0 gave you larger slenderness ratios and they still weren't greater than 40. Now you are going to say the tower columns weren't in the short column category and would have been subject to inelastic buckling. The AISC equations you show here and which you used in your paper are conservative for design.

You want to say the tower columns would fail due to buckling. Well how about a test case were an I beam with a slenderness ratio of 20 or lower failed due to inelastic buckling. Do you have any test cases? I have an AISC manual right here. I am familiar with the equations and monograph. You want to go around asking others if they have a clue and you seem to be the one who should be asked that question Mr. Smarty pants.

This is where Mr. Szamboti shows his lack of knowledge as regards to structural engineering. The slenderness ratio that we are talking about, and what Mr. Szamboti struggles to understand, here is defined as K*L/r.

Where:
k = effective length factor
L = length of the column (in)
r = radius of gyratio of the column (in) - [This is a function of the geometric properties of a column]

For the columns that we are talking about, the variables L and r are very well defined and not argued. The effective length factor 'k' is where he slips up. In the commentary of the AISC Manual of Steel Construction (arguably the Bible of how to design steel in structures) this factor 'k' is defined. The first place is in table C2.2 (shown below).

Table C-C2.2​

Under column (a) it defines the theoretical k value of 0.5 and a recommended design value of 0.65. It's fairly easy to see that this is where Mr. Szamboti thinks the factor k is defined, as the tower exterior columns were moment frames, which means that the top and bottom portions of the columns were fixed. Column (a) shows a column element fixed at the top and bottom, so he used it. And he's very wrong. The commentary clearly explains how this table is to be used on the page before the table, "These range from simple idealizations of single columns such as shown in Table C-C2.2 to complex buckling solutions for specific frames and loading conditions". In his rush to prove me wrong, I can only surmise that he went through the commentary to find an answer to his question, and stumbling upon the first table that seemed to show an answer that confirmed his bias, he lept to a conclusion. An incorrect one not supported by the document he was referencing.

The correct way to calculate this effective length factor is with the nomograph chart, shown below. The nomograph table is for frames which can translate horizontally, this contributes significantly to the stability of the frame.

Figure C-C2.4​

This table looks nonsensical, but it's fairly simple to use. First Ga and Gb need to be defined, which are simply a comparison of the stiffnesses of the columns to the girders. Ga is the comparative stiffness of the top point of the column and Gb is of the bottom. Then, to get the effective length factor, one merely needs to draw a straight line between these two points (see the figure below with Ga = 1.0 and Gb slightly stiffer).

Example​

In this example, the k factor of a frame that has a column stiffness that is roughly equal to the girder stiffness is about 1.4.

It is very easy to see with this table that the lowest factor k that a column in a moment frame can have is 1.0, rendering Mr. Szamboti's statement that it should be 0.5 or 0.65 completely without merit. The purpose of this isn't to belittle or attack Mr. Szamboti as being an incompetent engineer. I'm sure he is an excellent mechanical engineer, however he is not an expert in structural engineering, far from it. He is most definitely unqualified to "peer-review" papers of a structural engineering focus for anyone.

 

[jhunter1163]

Great post, NB. Even I, a non-engineer, could understand exactly why Tony the Twoofer is wrong.

I've also got one beverage of your choice that says he doesn't answer this criticism.

 

[uk_dave]

something technical with equations and stuff

[truther] And where on youtube can I view this? [/truther]

 

[A W Smith]

you left out one of the best parts

[realcddeal] Dont you mean Monograph? [/realcddeal]

 

[DGM]

It seems lately that they are just trying to sound smart. They don't seem willing to defend the articles just get them out. This has all the signs of a "recruit the uninformed" drive.
It doesn't matter to them if the information is correct as long as they have followers willing to repeat it.

 

[Newtons Bit]

Great post, NB. Even I, a non-engineer, could understand exactly why Tony the Twoofer is wrong.

I've also got one beverage of your choice that says he doesn't answer this criticism.

He's just quite simply wrong, and it's easy to see how. This is an example where the peer reviewers at J911 can be challenged on their expert status fairly easily. I just hope people like Lisabob are paying attention.

 

[Coffee]

He's just quite simply wrong, and it's easy to see how. This is an example where the peer reviewers at J911 can be challenged on their expert status fairly easily. I just hope people like Lisabob are paying attention.

Lisabob2 will ignore this. Lisabob2 routinely ignores posts that show the peer review at J911 to be a joke just as s/he ignores anything that go against the words of Jones and Griscom.

 

[Tony Szamboti]

I've never really been generally impressed by the folks at the "Journal" of 911 Studies, but my latest encounter with one of their peer reviewers, Tony Szamboti, on this forum leads me to believe that they're not only suffering from group think, but also egregiously incompetent on issues they proclaim to be experts on. Mr. Szamboti posited this question to a group of non-engineers:



I saw this question and jumped into the discussion, posting:


Mr. Szamboti's replied:



This is where Mr. Szamboti shows his lack of knowledge as regards to structural engineering. The slenderness ratio that we are talking about, and what Mr. Szamboti struggles to understand, here is defined as K*L/r.

Where:
k = effective length factor
L = length of the column (in)
r = radius of gyratio of the column (in) - [This is a function of the geometric properties of a column]

For the columns that we are talking about, the variables L and r are very well defined and not argued. The effective length factor 'k' is where he slips up. In the commentary of the AISC Manual of Steel Construction (arguably the Bible of how to design steel in structures) this factor 'k' is defined. The first place is in table C2.2 (shown below).

[qimg]http://www.internationalskeptics.com/forums/imagehosting/thum_16329477c5ad0df89f.jpg[/qimg]
Table C-C2.2​

Under column (a) it defines the theoretical k value of 0.5 and a recommended design value of 0.65. It's fairly easy to see that this is where Mr. Szamboti thinks the factor k is defined, as the tower exterior columns were moment frames, which means that the top and bottom portions of the columns were fixed. Column (a) shows a column element fixed at the top and bottom, so he used it. And he's very wrong. The commentary clearly explains how this table is to be used on the page before the table, "These range from simple idealizations of single columns such as shown in Table C-C2.2 to complex buckling solutions for specific frames and loading conditions". In his rush to prove me wrong, I can only surmise that he went through the commentary to find an answer to his question, and stumbling upon the first table that seemed to show an answer that confirmed his bias, he lept to a conclusion. An incorrect one not supported by the document he was referencing.

The correct way to calculate this effective length factor is with the nomograph chart, shown below. The nomograph table is for frames which can translate horizontally, this contributes significantly to the stability of the frame.

[qimg]http://www.internationalskeptics.com/forums/imagehosting/thum_16329477c57124fc1f.jpg[/qimg]
Figure C-C2.4​

This table looks nonsensical, but it's fairly simple to use. First Ga and Gb need to be defined, which are simply a comparison of the stiffnesses of the columns to the girders. Ga is the comparative stiffness of the top point of the column and Gb is of the bottom. Then, to get the effective length factor, one merely needs to draw a straight line between these two points (see the figure below with Ga = 1.0 and Gb slightly stiffer).

[qimg]http://www.internationalskeptics.com/forums/imagehosting/thum_16329477fcfe5ad3f4.jpg[/qimg]
Example​

In this example, the k factor of a frame that has a column stiffness that is roughly equal to the girder stiffness is about 1.4.

It is very easy to see with this table that the lowest factor k that a column in a moment frame can have is 1.0, rendering Mr. Szamboti's statement that it should be 0.5 or 0.65 completely without merit. The purpose of this isn't to belittle or attack Mr. Szamboti as being an incompetent engineer. I'm sure he is an excellent mechanical engineer, however he is not an expert in structural engineering, far from it. He is most definitely unqualified to "peer-review" papers of a structural engineering focus for anyone.

First, you are using the sideways uninhibited nomograph when you should be using the sideways inhibited nomograph. Paragraph 1.8.2 of the Commentary of the AISC manual says connections to floor slabs constitute a braced frame for horizontal stability. This would be considered sideways inhibited and there weren't any columns in the twin towers that did not have floor slabs at their beam connections. Here is a little tutorial on determining K factors for your prematurely cheering fans. Notice how those that are sideways inhibited are less than 1.0.

http://cnx.org/content/m10746/latest/

It sounds like you use the sideways uninhibited nomograph for conservativism.

Oh, and please spare us your canard about mechanical engineers not being structural engineers. What a load of baloney that is and you know it. The structural related courses for mechanical and civil engineers are the same in most schools. Where they diverge is that civils then do more with geotechnical (soils) and transportation (highway building) related courses and mechanicals do more with heat transfer and thermodynamic related courses in the other parts of their curriculums. Buckling can occur in machine elements just as it can with building frames. Mechanical engineers also design the structures for aircraft, automobiles, trains, etc. and buckling is certainly a potential failure mechanism there. There are also dynamic loads involved in these situations which is more complicated.

 

[Gravy]

Dude doesn't even know the difference between sideways and sidesway.

Which of the columns in your link http://cnx.org/content/m10746/latest/ best represents the exterior columns of the towers, Tony?

Mechanical engineers also design the structures for aircraft, automobiles, trains, etc. and buckling is certainly a potential failure mechanism there. There are also dynamic loads involved in these situations which is more complicated.

You work on antenna design, right?

 

[Tony Szamboti]

Dude doesn't even know the difference between sideways and sidesway.

Which of the columns in your link http://cnx.org/content/m10746/latest/ best represents the exterior columns of the towers, Tony?


You work on antenna design, right?

Mark, it sounds like the dumbed down version wasn't dumb enough for you.

How about column AB representing the perimeter columns. Do you get it? In fact since they would have been considered fixed on both ends the K factor would have been even lower than the pinned connection shown which has a K factor of .77.

Oh, and you caught a little spelling error. Congratulations. Are you going to try and declare victory with that? You can make a video of how many times you caught spelling errors of people you disagree with.

How did the core columns fail in the towers and what was their K factor? It wasn't Newtons Bit's 1.0 as even the previously blind cheering fans here at JREF must now admit.

 

[T.A.M.]

I'll soon be expecting someone to stick out their tongue and shout "Nah Nah Nah Nah Nah Nah!"

TAM

 

[Myriad]

First, you are using the sideways uninhibited nomograph when you should be using the sideways inhibited nomograph. Paragraph 1.8.2 of the Commentary of the AISC manual says connections to floor slabs constitute a braced frame for horizontal stability. This would be considered sideways inhibited and there weren't any columns in the twin towers that did not have floor slabs at their beam connections. Here is a little tutorial on determining K factors for your prematurely cheering fans. Notice how those that are sideways inhibited are less than 1.0.

http://cnx.org/content/m10746/latest/

It sounds like you use the sideways uninhibited nomograph for conservativism.

The exercise you reference specifically states that the sidesway inhibited condition of AB, CD, and GF is due to the support at point J. The little triangular thingie in the schematic at point J is a symbol representing a strong immovable anchor point, such as a concrete foundation pier, correct?

What would be the equivalent of point J on the above-ground floors of the World Trade Center?

Seems to me (with all due respect for your expertise) that a WTC tower floor would be more like EH, and the columns more like DE and GH, which are described as sidesway uninhibited in the exercise you referenced. That would agreee with Newton's assessment.

Respectfully,
Myriad

 

[Tony Szamboti]

The exercise you reference specifically states that the sidesway inhibited condition of AB, CD, and GF is due to the support at point J. The little triangular thingie in the schematic at point J is a symbol representing a strong immovable anchor point, such as a concrete foundation pier, correct?

What would be the equivalent of point J on the above-ground floors of the World Trade Center?

Seems to me (with all due respect for your expertise) that a WTC tower floor would be more like EH, and the columns more like DE and GH, which are described as sidesway uninhibited in the exercise you referenced. That would agreee with Newton's assessment.

Respectfully,
Myriad

You can argue with the AISC manual, which says floor slab connections constitute a braced frame. The reason is obvious and it is the stiffness and great inertia which the floor slabs impart to the connection. Newtons Bit's use of a K factor of 1.0 is conservative for design, as I pointed out earlier, and does not belong in determining what the real K factor was in the failure analysis.

 

[peteweaver]

Just because the structure below the impact areas of the WTC's were untouched by fire, and undamaged before the collapse, it doesn't mean they were indestructible.
They were capable of supporting the mass of the structure above them, yes, but not, when that mass became dynamic, and fell.

Velocity changes things somewhat.

 

[Tony Szamboti]

Just because the structure below the impact areas of the WTC's were untouched by fire, and undamaged before the collapse, it doesn't mean they were indestructible.
They were capable of supporting the mass of the structure above them, yes, but not, when that mass became dynamic, and fell.

Velocity changes things somewhat.

You need to show how you get to the dynamic load. I am very skeptical that all of the columns on the initiation floors could have buckled simultaneously, due to fire, and allowed a freefall to create any dynamic load. This is why we are discussing the possibility of buckling and the actual slenderness ratio of the columns plays a big part in that possibility or lack of.

With Gregory Urich's substantiated mass analysis it turns out that the factor of safety of the central core columns was approximately 3.00 to 1 and the perimeter columns were a minimum of 5.00 to 1, when considering gravity loads only. This would need to be overcome to allow a collapse.

 

[Tony Szamboti]

I'll soon be expecting someone to stick out their tongue and shout "Nah Nah Nah Nah Nah Nah!"

TAM

No, that won't happen. However, I would appreciate an apology for being called a moron by a prominent poster here.

 

[Myriad]

You can argue with the AISC manual, which says floor slab connections constitute a braced frame. The reason is obvious and it is the stiffness and great inertia which the floor slabs impart to the connection. Newtons Bit's use of a K factor of 1.0 is conservative for design, as I pointed out earlier, and does not belong in determining what the real K factor was in the failure analysis.

My copy of the AISC manual (Specification for Structural Steel Buildings, March 9, 2005) does not contain any paragraph 1.8.2, nor does it use the phrase "floor slab" except in only one place, in a section concerned with fire resistance. Thus, I cannot verify your reference to what the AISC manual says about floor slab connections. Please be more specific in the reference (chapter and section name) or quote a sentence or two directly so that I can find the equivalent passage the current manual.

Without a specific valid reference, I can't tell whether the statement you are referencing is correctly interpreted as applying to the types of floors in the WTC.

It seems unlikely that the connections between the columns and floors in the WTC fit the definition of sidesway inhibited, because this would require them (according to table C-C2.2) to be "rotation fixed" when clearly in structural terms they are hinged connections. The floor "slabs" in the WTC towers are not thick enough to provide significant resistance to the rotation of a column connection, even if the columns had been embedded in the concrete rather than, as they actually were, bolted to the ends of the floor trusses.

Furthermore, the towers as a whole seem to far better fit AISC's definition of a moment frame than a braced frame. They did not resemble in any way a "vertical truss system" due to the almost complete lack of diagonal bracing members.

So again, with the exception of the one reference to Commentary paragraph "1.8.2" discussing "floor slabs" that you've offered, neither of which are found in a search of the Commentary, all of the AISC documentation seems to support Newton's view.

Respectfully,
Myriad

 

[Tony Szamboti]

My copy of the AISC manual (Specification for Structural Steel Buildings, March 9, 2005) does not contain any paragraph 1.8.2, nor does it use the phrase "floor slab" except in only one place, in a section concerned with fire resistance. Thus, I cannot verify your reference to what the AISC manual says about floor slab connections. Please be more specific in the reference (chapter and section name) or quote a sentence or two directly so that I can find the equivalent passage the current manual.

Without a specific valid reference, I can't tell whether the statement you are referencing is correctly interpreted as applying to the types of floors in the WTC.

It seems unlikely that the connections between the columns and floors in the WTC fit the definition of sidesway inhibited, because this would require them (according to table C-C2.2) to be "rotation fixed" when clearly in structural terms they are hinged connections. The floor "slabs" in the WTC towers are not thick enough to provide significant resistance to the rotation of a column connection, even if the columns had been embedded in the concrete rather than, as they actually were, bolted to the ends of the floor trusses.

Furthermore, the towers as a whole seem to far better fit AISC's definition of a moment frame than a braced frame. They did not resemble in any way a "vertical truss system" due to the almost complete lack of diagonal bracing members.

So again, with the exception of the one reference to Commentary paragraph "1.8.2" discussing "floor slabs" that you've offered, neither of which are found in a search of the Commentary, all of the AISC documentation seems to support Newton's view.

Respectfully,
Myriad

I have the Eighth edition of the AISC manual and the Chapter 5 Commentary was effective 11/1/1978. Maybe they changed the paragraph number in later editions. Chapter 5 Section 1.8 of the Eighth edition is titled "Stability and Slenderness Ratios". You should look for a paragraph with that title.

I think you need to do an analysis to show that any of the columns in the towers would have performed as pinned connections rather than fixed on both ends. Diagonal bracing is not the only thing that causes a frame to act as braced.

 

[rwguinn]

I have the Eighth edition of the AISC manual and the Chapter 5 Commentary was effective 11/1/1978. Maybe they changed the paragraph number in later editions. Chapter 5 Section 1.8 of the Eighth edition is titled "Stability and Slenderness Ratios". You should look for a paragraph with that title.

I think you need to do an analysis to show that any of the columns in the towers would have performed as pinned connections rather than fixed on both ends. Diagonal bracing is not the only thing that causes a frame to act as braced.

I do not have access right now to the manual, and seldom if ever use it in my job.
You are right that diagonal bracing is not the only thing that will cause it to act as braced--and the intact structure would act that way--outside panels acting as shear panels between verticals.
But the tower was not intact. The bracing (shear panels) had been compromised--breached, even.
The floors do not restrain the verticals from rotation, however, and the top of a set of columns is assuredly not fixed. The whole floor is able to translate WRT the floor below, especially when the shear panels go away.

Also you are using "conservative" in a way I am unfamiliar with.
A conservative design, to me, means designed to the worst-possible scenario--and a conservative analysis means the worst possible load case with the worst possible boundary conditions--or the most likely to cause failure. You seem to be implying it is something opposite that...

 

[Newtons Bit]

First, you are using the sideways uninhibited nomograph when you should be using the sideways inhibited nomograph. Paragraph 1.8.2 of the Commentary of the AISC manual says connections to floor slabs constitute a braced frame for horizontal stability. This would be considered sideways inhibited and there weren't any columns in the twin towers that did not have floor slabs at their beam connections. Here is a little tutorial on determining K factors for your prematurely cheering fans. Notice how those that are sideways inhibited are less than 1.0.

http://cnx.org/content/m10746/latest/

It sounds like you use the sideways uninhibited nomograph for conservativism.

Oh, and please spare us your canard about mechanical engineers not being structural engineers. What a load of baloney that is and you know it. The structural related courses for mechanical and civil engineers are the same in most schools. Where they diverge is that civils then do more with geotechnical (soils) and transportation (highway building) related courses and mechanicals do more with heat transfer and thermodynamic related courses in the other parts of their curriculums. Buckling can occur in machine elements just as it can with building frames. Mechanical engineers also design the structures for aircraft, automobiles, trains, etc. and buckling is certainly a potential failure mechanism there. There are also dynamic loads involved in these situations which is more complicated.

First, download and read the commentary for AISC-360-05. At least this way we can be referencing the same document. Then, admit you've made a mistake because you've yet again failed completely to understand what you're talking about. It's a simple thing to do.

Let's look at the example you've provided. The author provides a diagram representing a building section:



The important thing to note is that point J is a pinned connection. The author is correct in saying that sidesway is inhibited because point J, and thus the beams connected to it, cannot translate. Point J represents a stiff lateral element, a braced frame or a shear wall perhaps. It does not represent a moment frame.

The author then shows how to calculate k for the top portion of the building, columns DE and GH. These he says are, "there is no sideways bracing for the top portion of the frame". That's because they are not connected to a braced frame or a shear wall but rather a moment frame (or a cantilevered column, it depends on what the connections look like).